Experimental Program

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Experimental Program

This chapter discusses several key issues of the specimen design: choice of dimensions, detailing of the PR connections, design of the interface headed studs and design of wall reinforcement. The test setup is illustrated and material properties of the specimen are reported. The instrumentation plan is then specified to capture both the global and local behavior of the specimen.
3.1 Specimen design

3.1.1 Specimen dimensions

The two-story, one bay specimen, designed according to the similitude requirements (Sabnis et al., 1983) that relate the model to the prototype, was meant to represent the bottom two levels of the six story prototype infilled steel frame at approximate one-third scale. Figure 2.3.1 shows the dimension and member size for the six-story prototype structure. To scale down the prototype structure to the specimen, two independent scale factors, S and Sl,, were chosen for stress  and length l, respectively; all remaining scale factors were either equal to unity or were functions of S and Sl. For example, the scale factor of moment of inertia was Sl4 and that of the section area was Sl2. The resulting specimen size after scaling is shown in Figure 3.1.1. The steel columns

comprised W5x19 wide-flange shapes, while the steel beams comprised W8x13 wide-flange shapes. The column material was A572 Grade 50 steel and the beam material was

A36 steel, the same as those of the prototype structure. The RC infill wall was 3.5 inch thick, with the 28 day compressive strength of the concrete targeted at 3.5 ksi. Number 2 deformed reinforcing steel bars were used as the infill wall reinforcement. These bars were purchased from CTL inc. and originated from a mill in Switzerland. They have similar material properties to domestic reinforcing bars in regular sizes. Partially-restrained connections were used to connect the middle beam and the top beam to the columns. Table 3.1.1 compares the geometry and material-related properties of the prototype structure and the specimen. For a practical true model involving reinforced concrete (Sabnis et al., 1983), S shall be 1 and Sl shall be 3 in one-third scaling in order

Table 3.1.1 Geometry and Material Properties

of the Prototype Structure and the Specimen

Item ID


Prototype structure


Scale factor

Ideal Scale factor


Story height (inches)





Frame spacing (inches)





Wall thickness (inches)





Section area of the column (inch2)





Web area of the column (inch2)





Moment of inertia of the column (inch4)





Plastic modulus of the column (inch3)





Section area of the beam (inch2)





Web area of the beam (inch2)





Moment of inertia of the beam (inch4)





Plastic modulus of the beam (inch3)






Nominal yield strength of frame steel (ksi)





Elastic modulus frame steel (ksi)





Nominal compressive strength of wall

concrete (ksi)





Yield strength of reinforcing bar (ksi)





Elastic modulus of concrete (ksi)





to meet the similitude requirement. As a result, the ideal scale factors for geometry and material properties were determined in accordance to their dimensions and are listed in Table 3.1.1. Due to the limited sizes of commercially available wide-flange shapes, the

W519 column is the section that is both compact and has the closest scale factors to the ideal scale factors for its geometrical properties. The scale factors for the geometrical properties of the W813 beam are approximately 70%-75% of the ideal scale factors. However, the numerical analysis of the prototype structure indicated the beam size had little effect on the structural behavior of this composite system during the elastic range. Furthermore, according to Liauw and Kwan (1983a, 1983b), it was the moment capacity of the connection that played an important role in determining the failure modes and maximum lateral strength of this type of composite system. Therefore, it was believed that a little oversize of the beam would not affect the structural behavior of the specimen as long as the partially restrained connection was properly detailed. The item that has the maximum discrepancy between the scale factor and the ideal factor is the frame spacing, which was scaled down 4.18 times from the prototype structure to the specimen. The major reason for adopting this value was to insure that the maximum capacity of the loading system, which was provided by two 110 kips actuators, was sufficient. According to the equivalent lateral force procedure in NEHRP (1997), the design base shear force for the six-story prototype structure was 948 kips. This lateral force was then distributed vertically along the structure to compute the structural response and select the appropriate members. It is of great interest to know the specimen responses under the corresponding scaled design lateral force. Because the majority of the design lateral force is to be carried by the RC infill wall, the approximate scale factor of the design lateral force could be estimated in accordance with the cross-section dimensions of the infill wall and concrete properties. Dimensional analysis of the reinforced concrete shows that the scale factor of a concentrated load is S(Sl)2. Comparison of the prototype structure and the specimen resulted in:

Therefore, the approximate scale factor for the design lateral force was 13.3, which yielded a design base shear force of the specimen as approximately 71 kips. This design lateral force was used in detailing the headed stud connectors along the interfaces and the reinforcement in the infill wall, as shown in the following sections.

3.1.2 Beam-to-Column Connection Design

In almost all of the previous static or cyclic tests on infilled steel frames, the beams were welded to the columns across the full cross section so as to act as a fully-restrained (FR) connection (Mallick and Severn, 1968; Liauw and Kwan, 1983a, 1983b; Makino, 1984; Kwan and Xia, 1995). Recently, Nadjai and Kirby (1998) analyzed the behavior of non-composite infilled steel frames with semi-rigid connections using the finite element method. In the specimen tested in this work, PR connections comprising a top-and-seat-angle and a double-web angle were used for the following reasons. First, in the steel frame with composite reinforced concrete infill wall system, the girders are mainly used to transfer the lateral shear forces and carry gravity loads. The moment-rotation response of the girders has little effect on the system behavior in the elastic stages of lateral loading (Nadjai and Kirby, 1998). Finite element analysis of the composite reinforced concrete infilled frames indicated that there was no major difference between the elastic behavior of the system with FR connections and that using PR or pin connections (Tong et al., 1998). Therefore, expensive FR connections are not warranted at low load levels. Second, with an increase of lateral load and crushing of the concrete panel corners, the girders will be required to resist a certain amount of moment, which is induced by the spread of the corner interface normal forces caused by the concrete crushing. Also, with the degradation of the concrete panel stiffness, the steel frame must contribute its lateral stiffness to ensure the integrity of the system later in an earthquake. Thus, some amount of connection rotational stiffness, strength, and ductility are required. Third, the failure mechanism and a corresponding design procedure of the PR connection comprising a top-and-seat-angle and a double-web angle have been provided by Kim and Chen (1998). Although their design formulas are based on the study of static behavior of bare steel frames, this type of PR connection has proven to have stable hysteretic behavior when subjected to cyclic load (Leon and Shin, 1995).

Figure 3.1.2 provides the detailing of the PR connection in this specimen. The deflection pattern of this PR connection subjected to bending is shown in Figure 3.1.3 (Kim and Chen, 1998). It can be seen that the collapse mechanism of the top angle may be modeled by the development of two plastic hinges, one at the edge of the bolt head and the other along the k-line of the vertical leg. The deformation pattern also shows that the web angles can contribute to the moment resisting capacity. Design of this PR connection is documented in Appendix A, which is adapted in part from the procedures proposed by Kim and Chen (1998). All the shear force was assumed to be carried by a L225/165 double-web angle. Because the PR connection may endure a high concentrated shear force caused by the compressive strut action of the concrete infill wall, the double-web angle and the corresponding high strength bolts were sized to have at least the same shear resistant capacity Vn as that of the W813 beam, thus reducing the possibility of a shear failure of the PR connection. This decision was made based on the fact that a moment-resistant failure mode of the connection usually has a better energy dissipation capacity than the shear failure mode. In the specimen, the double-web angle was connected to the column flange by using 1/2 inch A325 bolts, but had to be welded to the beam web because there was not enough clearance for bolting. Typically, the angles would be bolted to the beam web in a full-scale structure. However, this detailing will not change the failure mechanism of the web a ngles appreciably because the plastic hinges were expected to form in the angle legs that were bolted to the column flanges (Figure 3.1.3).

A L535/165 angle was chosen for the top and seat angles. The angle was selected so that the stiffness of the top or seat angle was less than that of the column flange, which is 0.43 inch thick. Each top and seat angle was connected to the beam

Fig. 3.1.3 Deformation of the PR Connection Subjected to Bending

[after Kim and Chen (1998)]

flange by using four 5/8 inch A490 bolts and was connected to the column flange by using two 5/8 inch A490 bolts. All bolts were detailed as slip critical. The ultimate moment capacity contributed by all parts of the connection was approximately 460 kip-inches, 88% of nominal plastic moment strength of the beam (Appendix A). However, corner crushing of the infill wall will induce a non-uniform bearing force on the top portion of the windward column or the bottom portion of the leeward column, especially on the portion near the beam-to-column joint. This bearing force may then induce a certain amount of tensile force in the PR connection. As a result, the PR connection is actually subjected to the interaction of moment and tension and the actual moment capacity of the PR connection will be reduced by the effect of the axial tensile force. The model of Kim and Chen (1998) only predicts the pure moment capacity of the PR connection and gives no consideration on the effect of an axial tensile force. In designing the PR connection of this specimen, the axial tensile force was assumed to equal the nominal shear resistance capacity of the column. Conservatively, a linear moment-axial force interaction equation was used to include the effect of the axial tensile force on the moment capacity of the PR connection. As calculated in Appendix A, the moment capacity of the PR connection was estimated to be approximately 50% of the nominal plastic moment strength of the beam due to the inclusion of the axial force.
3.1.3 Design of the Headed Stud Connectors along the Interfaces

Headed stud connectors were installed along the interfaces of the RC infill wall and the steel members to ensure the composite action of the specimen. The interface connectors should be designed in accordance with their two primary properties: strength and ductility. According to the possible plastic failure mechanisms of infilled steel frames (Liauw and Kwan, 1983a, 1983b), the strength of interface connectors has significant effect on the maximum strength of the entire composite system. This is discussed further in Chapter 9. Ductility of the interface connectors plays a major role in retaining system integrity and increasing energy dissipation capacity.

For a single headed stud connector loaded in shear in the concrete, there are four primary failure modes, as shown in Figure 3.1.4. Breaking-out of the concrete (Figure 3.1.4.(c)) usually occurs at connectors located near the free edges of the concrete. Prying-out of the concrete (Figure 3.1.4.(d)) only occurs to connectors with a small ratio of embedment depth to shank diameter. Using sufficient embedment depth usually can eliminate this type of failure. For the majority of the headed stud connectors loaded in shear, shearing of the connector shank or the crushing of the concrete in the bearing zone are the two significant failure modes, which were observed at the same time in the same specimen in many tests (Ollgaard et al., 1971). The shear strength of a single headed stud in AISC (1993) and PCI (1992) is determined in accordance with these two failure modes. However, as shown in Figure 3.1.5, due to the limited thickness of the infill wall, the dispersal of the concentrated shear load into the concrete may induce cracks in three different orientations in the concrete: ripping, shear, and splitting (Oehlers, 1989). The cracking tends to propagate into the bearing zone and relieve the tri-axial stress-field on this part of concrete, so that the concrete in the bearing zone will crush prematurely. As a result, the specified shear strength in standard codes can not be achieved and the ductility of the connectors is also reduced. For a group of headed stud connectors loaded in shear in the infill wall, Figure 3.1.5 shows that another possible

Fig. 3.1.4 Failure Modes of a Headed Stud Connector Loaded in Shear

(a) crushing of the concrete (b) shearing off of the connector shank

(c) breaking out of the concrete (d) prying out of the concrete

Fig. 3.1.5 Shear Force Transfer Mechanism of a Headed Stud Connector

[after Oehlers (1989) with minor changes]

failure mode is caused by cracking just above the head of the headed studs and through the entire section of the concrete.

Additionally, linear elastic finite element analysis of the prototype structure (Tong et al., 1999) indicated that the headed studs in this composite structural system, particularly in the corners of the panels, are under the interaction of tension and shear. This phenomenon was also observed in Liauw’s tests (Liauw and Kwan, 1983b, 1985). Therefore, the connectors placed at the corners are required to provide enough tensile strength and ductility to delay the separation between the RC infills and the steel frame members. In Liauw’s tests (Liauw and Kwan, 1983b, 1985), because of the weak confinement provided by the limited wall thickness, the maximum tensile strength of the headed stud connectors in the infill wall was controlled by concrete failure, involving the pull-out of a concrete cone. This type of connector failure generally gives relatively lower tensile strength and ductility, and thus the composite action may be deteriorated by the premature separation between the RC infills and the steel frame members.

In order to characterize these complex shear and failure modes, in an earlier portion of this research, a series of twelve experiments were conducted to investigate the strength and ductility of shear studs in RC wall panel (Saari , 1998). The parameters included: 1) monotonic or cyclic shear loading; 2) zero tensile loading or monotonic tensile loading with the force equal to approximately 50% of the nominal tensile strength of the headed stud; 3) confining with either perimeter bars placed along the base of the studs or with steel reinforcing cages (see Figure 3.1.6) and; 4) use of a ductility enhancing polymer cone. The tests showed that the reinforcing cage mitigated the primary types of cracking so that the fracture of the stud base in the base metal became the controlling failure mode. The impact of using the reinforcing cage resulted in the increase of stud strength and ductility. When axial tension, approximately 50% of the nominal tensile strength of the stud, was applied, the monotonic shear strength was 27% larger when the confining cage was used instead of the perimeter bar scheme, and the maximum slip achieved was three times larger. The impact of applying the axial tension on the stud resulted in the decrease of the shear strength. In the monotonic tests on the specimens with confining cages, the

(a) Reinforcing cage (b) Perimeter bar

Fig. 3.1.6 Confinement of Headed Stud Connectors
shear strength of the stud reduced by nearly 40% when the same amount of axial tension was applied. The introduction of cyclic loading translated to a loss of approximately 17% in the stud shear strength, and resulted in a reduction of 70% to 80% in the deformation capacity displayed by the specimen with confining cages. A ductility enhancing device, comprising a plastic cone placed around the shank at the base of the stud, increased deformation capacity by approximately 60% for cyclic loading.

Because adequate confining reinforcement is critical to achieve strength and ductility for studs in an infill wall, steel confining cages were adopted in this specimen. The headed stud was 3/8 inches in diameter and 2.5 inches in total length, so that the ratio of stud length to diameter (2.5 to 3/8) was the same as that of a typical full scale stud (5 to 3/4). The thickness of the head itself was 5/16 inches and the diameter of the head was 3/4 inches.

The design shear strength of the stud was determined based on the following equation:



c = compressive strength of the concrete of the infill wall, ksi

c = young’s modulus of the concrete of the infill wall, ksi

u = minimum specified tensile strength of the stud material, ksi

Asc = cross section area of the shear stud shaft, inch2

Equation (3.1.1) was modified from AISC (1993) with one additional factor 1 to represent the detrimental effect of low cycle fatigue on the shear stud connectors during an earthquake. The 15% strength reduction resulting from using 1 was based on the cyclic stud tests conducted as part of this research project (Saari, 1998).

Due to the presence of confining cages, the design tensile strength of the 3/8 inch stud was governed by the stud shank failure and determined in accordance to the following formula provided by PCI (PCI, 1992):



Fy = nominal yield strength of a stud connector, ksi

The effect of the cyclic loading on the tensile strength of the stud was not considered due to the lack of sufficient experimental results.

For headed studs subjected to combined cyclic shear and tension in an infill wall, the strength capacity was governed by the following equation from PCI (PCI, 1992):



 = 1.0

As seen from Eq. (3.1.3), the presence of the tensile force in the stud can result in a significant reduction of shear strength of the headed stud. If the tensile force is 50% of the tensile strength of the stud, the maximum allowable shear strength will be reduced by 14%. In order to include the effect of axial force in design, it was assumed conservatively that the shear strength of a stud should be neglected as long as Eq. (3.1.3) was breached based on the elastic analysis of the structure at the design force level.

The finite element analysis results of the prototype structure were used to estimate the percentage of the studs that breached the interaction Eq. (3.1.3). For a headed stud used in the prototype structure with sufficient confinement (the stud was 3/4 inches in diameter and 5 inches in length), Eq. (3.1.1) gives a design shear strength of approximately 22 kips and Eq. (3.1.2) gives a design tensile strength of approximately 24 kips. The internal forces of the studs along the most critical interface, the bottom interface of the first story, are listed in Table 2.3.2 due to the lateral load combination discussed in Chapter 2. It can found that approximately 25% of the studs (11 studs from the total of 41 studs) along the bottom interface breach the interaction Eq. (3.1.3).

It was assumed that, as in the prototype structure, 25% of the studs along the beam-infill interface in the specimen would breach the interaction equation at the design force level, and the remaining 75% of the studs needed to resist 100% of the design lateral load. As discussed in Section 3.1.1, the design lateral load for the specimen was determined to be approximately 71 kips. As a result, 18 headed studs were placed along the beam-infill wall interface at 4 inch spacing to transfer the 71 kips design lateral load (75% of the total shear capacity of 18 studs is 69 kips, which is smaller than the 71 kips design lateral load by 2.8%). It was also assumed that the shear force per unit length along the column–infill interface was the same as that along the beam–infill interface. Therefore, the headed studs along the column–infill interface were also placed at 4 inch spacing.

The confining reinforcement cages, comprising rectangle hoops and four horizontal bars, were placed all around the infill wall perimeter in the specimen. The detailing of confining cages is shown in Figures 3.1.6 and 3.1.7. The rectangle hoop was 5 inches high and 2.5 inches wide, arranged at 2 inch spacing. Four horizontal bars were tied to the corners of the rectangle hoop to form the confining cage. Cold-drawn No.3 gage steel was used to construct the rectangle hoops and No. 2 reinforcing bar was used as the horizontal bars. These sizes hold to the one-third scale requirements relative to the dimensions of confining cages that would be used in a full-scale infill wall (Saari, 1998).

3.1.4 Design of the Wall Reinforcement

The infill wall of the specimen was assumed to carry 100% of the lateral design shear force and 40% of the corresponding overturning moment. According to Section in the ACI building code (ACI, 1995), a reduction factor of 0.6 was used to calculate the shear strength of the RC infill walls. Both the horizontal and the vertical reinforcement ratios were 0.51%, achieved by two curtains of No. 2 bars spaced at 5.5

inches. The estimated shear strength of the RC infill wall was 78.5 kips. Detailing of the wall reinforcement is shown in Figure 3.1.7.

A key problem of is to develop the reinforcing bars. The development length of a No. 2 bar is approximately 6 inches without a hook at the end or 5.1 inches with a hook in accordance with ACI (1995). Therefore, the height of the reinforcement cage theoretically should be at least great than 5.1 inches to ensure that the reinforcement can approach its yield strength within the confining reinforcement cage. In actual construction, No. 6 or No. 7 bars are usually used as the wall reinforcement. If the concrete compressive strength of the prototype structure is 4000 psi and a hook is used at the bar end, the development length will be approximately 14.2 inches for No. 6 bars and 16.6 inches for No. 7 bars. That means the reinforcement cage in the prototype structure should be at least 14-15 inches high. Such a reinforcement cage is feasible but may be not economical in practice. However, if the wall reinforcement has the same enlarged end as the stud head, a portion of the tensile force in the connectors may be transferred through direct strut action between the reinforcement head and the surrounding concrete so that the development length can be reduced. Such reinforcement is called T-headed reinforcement and is manufactured by welding a steel plate head to the end of a standard reinforcing bar or by threading a nut to the end of a bar. An increasing amount of research work on T-headed reinforcement has been done in recent years because of its advantages in detailing, construction, and fabrication (DeVries, 1996; Wallace et al., 1998). It has been shown that a development length of 8 to 10 bar diameters is sufficient to develop the yield stress of a bar having a T-head (Bode and Roik, 1987; Balogh et al., 1991). For the above reasons, T-headed reinforcing bars were used in the present project, which was fabricated by electric arc welding a 3/8 inch nut to the end of the No.2 bar.

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